Design of Steel Structures

Prof. S.R.Satish Kumar and Prof. A.R.Santha Kumar

6.3 Behaviour of steel beams
Laterally stable steel beams can fail only by (a) Flexure (b) Shear or (c) Bearing,
assuming the local buckling of slender components does not occur. These three
conditions are the criteria for limit state design of steel beams. Steel beams would also
become unserviceable due to excessive deflection and this is classified as a limit state
of serviceability.

The factored design moment, M at any section, in a beam due to external actions
shall satisfy
M ≤ Md
Where Md = design bending strength of the section

6.3.1 Design strength in bending (Flexure)
The behaviour of members subjected to bending demonstrated in Fig 6.1

Fig 6.1 Beam buckling behaviour

This behaviour can be classified under two parts:

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Design of Steel Structures

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• When the beam is adequately supported against lateral buckling, the beam
failure occurs by yielding of the material at the point of maximum moment. The beam is
thus capable of reaching its plastic moment capacity under the applied loads. Thus the
design strength is governed by yield stress and the beam is classified as laterally
supported beam.

• Beams have much greater strength and stiffness while bending about the
major axis. Unless they are braced against lateral deflection and twisting, they are
vulnerable to failure by lateral torsional buckling prior to the attainment of their full inplane plastic moment capacity. Such beams are classified as laterally supported beam.

Beams which fail by flexual yielding
Type1: Those which are laterally supported
The design bending strength of beams, adequately supported against buckling
(laterally supported beams) is governed by yielding. The bending strength of a laterally
braced compact section is the plastic moment Mp. If the shape has a large shape factor
(ratio of plastic moment to the moment corresponding to the onset of yielding at the
extreme fiber), significant inelastic deformation may occur at service load, if the section
is permitted to reach Mp at factored load. The limit of 1.5My at factored load will control
the amount of inelastic deformation for sections with shape factors greater than 1.5.
This provision is not intended to limit the plastic moment of a hybrid section with a web
yield stress lower than the flange yield stress. Yielding in the web does not result in
significant inelastic deformations.

Type2: Those which are laterally shift
Lateral-torsional buckling cannot occur, if the moment of inertia about the
bending axis is equal to or less than the moment of inertia out of plane. Thus, for
shapes bent about the minor axis and shapes with Iz = Iy such as square or circular

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Design of Steel Structures

Prof. S.R.Satish Kumar and Prof. A.R.Santha Kumar

shapes, the limit state of lateral-torsional buckling is not applicable and yielding controls
provided the section is compact.

6.3.1.1 Laterally supported beam
When the lateral support to the compression flange is adequate, the lateral
buckling of the beam is prevented and the section flexual strength of the beam can be
developed. The strength of I-sections depends upon the width to thickness ratio of the
compression flange. When the width to thickness ratio is sufficiently small, the beam
can be fully plastified and reach the plastic moment, such section are classified as
compact sections. Howeverprovided the section can also sustain the moment during the
additional plastic hinge rotation till the failure mechanism is formed. Such sections are
referred to as plastic sections. When the compression flange width to thickness ratio is
larger, the compression flange may buckle locally before the complete plastification of
the section occurs and the plastic moment is reached. Such sections are referred to as
non-compact sections. When the width to thickness ratio of the compression flange is
sufficiently large, local buckling of compression flange may occur even before extreme
fibre yields. Such sections are referred to as slender sections.
The flexural behaviour of such beams is presented in Fig. 6.2. The section
classified as slender cannot attain the first yield moment, because of a premature local
buckling of the web or flange. The next curve represents the beam classified as 'semicompact' in which, extreme fibre stress in the beam attains yield stress but the beam
may fail by local buckling before further plastic redistribution of stress can take place
towards the neutral axis of the beam. The factored design moment is calculated as per
Section 8.2 of the code.

The curve shown as 'compact beam' in which the entire section, both
compression and tension portion of the beam, attains yield stress. Because of this
plastic redistribution of stress, the member o attains its plastic moment capacity (Mp) but
fails by local buckling before developing plastic mechanism by sufficient plastic hinge

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rotation. The moment capacity of such a section can be calculated by provisions given
in Section 8.2.1.2. This provision is for the moment capacity with low shear load.

Fig 6.2 Flexual member performance using section classification
Low shear load is referred to the factored design shear force that does not
exceed 0.6Vd, where Vd is the design shear strength of cross section as explained in
8.2.1.2 of the code.

Fig.6.3 Interaction of high shear and bending moment

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Design of Steel Structures

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6.3.1.1.1 Holes in the tension zone
The fastener holes in the tension flange need not be allowed for provided that for
the tension flange the condition as given in 8.2.1.4 of the code is satisfied. The
presence of holes in the tension flange of a beam due to connections may lead to
reduction in the bending capacity of the beam.

6.3.1.1.2 Shear lag effects
The simple theory of bending is based on the assumption that plane sections
remain plane after bending. But, the presence of shear strains causes the section to
warp. Its effect in the flanges is to modify the bending stresses obtained by the simple
theory, producing higher stresses near the junction of a web and lower stresses at
points away from it (Fig. 6.4). This effect is called ‘shear lag’. This effect is minimal in
rolled sections, which have narrow and thick flanges and more pronounced in plate
girders or box sections having wide thin flanges when they are subjected to high shear
forces, especially in the vicinity of concentrated loads. The provision with regard to
shear lag effects is given in 8.2.1.5.

Fig.6.4 Shear Lag effects

6.3.2 Laterally unsupported beams
Under increasing transverse loads, a beam should attain its full plastic moment
capacity. This type of behaviour in a laterally supported beam has been covered in

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Design of Steel Structures

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Section 8.2.1. Two important assumptions have been made therein to achieve the ideal
beam behaviour.
They are:
•

The compression flange of the beam is restrained from moving laterally;

•

Any form of local buckling is prevented

and

A beam experiencing bending about major axis and its compression flange not
restrained against buckling may not attain its material capacity. If the laterally
unrestrained length of the compression flange of the beam is relatively long then a
phenomenon known as lateral buckling or lateral torsional buckling of the beam may
take place and the beam would fail well before it can attain its full moment capacity.
This phenomenon has close similarity with the Euler buckling of columns triggering
collapse before attaining its squash load (full compressive yield load).

6.3.2.1 Lateral-torsional buckling of beams
Lateral-torsional buckling is a limit-state of structural usefulness where the
deformation of a beam changes from predominantly in-plane deflection to a combination
of lateral deflection and twisting while the load capacity remains first constant, before
dropping off due to large deflections. The analytical aspects of determining the lateraltorsional buckling strength are quite complex, and close form solutions exist only for the
simplest cases.

The various factors affecting the lateral-torsional buckling strength are:
•

Distance between lateral supports to the compression flange.

•

Restraints at the ends and at intermediate support locations (boundary

conditions).
•

Type and position of the loads.

•

Moment gradient along the length.

•

Type of cross-section.

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•

Non-prismatic nature of the member.

•

Material properties.

•

Magnitude and distribution of residual stresses.

•

Initial imperfections of geometry and loading.

They are discussed here briefly:
The distance between lateral braces has considerable influence on the lateral
torsional buckling of the beams.

The restraints such as warping restraint, twisting restraint, and lateral deflection
restraint tend to increase the load carrying capacity.

If concentrated loads are present in between lateral restraints, they affect the
load carrying capacity. If this concentrated load applications point is above shear centre
of the cross-section, then it has a destabilizing effect. On the other hand, if it is below
shear centre, then it has stabilizing effect.

For a beam with a particular maximum moment-if the variation of this moment is
non-uniform along the length (Fig. 6.5) the load carrying capacity is more than the beam
with same maximum moment uniform along its length.

If the section is symmetric only about the weak axis (bending plane), its load
carrying capacity is less than doubly symmetric sections. For doubly symmetric
sections, the torque-component due to compressive stresses exactly balances that due
to the tensile stresses. However, in a mono-symmetric beam there is an imbalance and
the resistant torque causes a change in the effective torsional stiffeners, because the
shear centre and centroid are not in one horizontal plane. This is known as "Wagner
Effect".

Indian Institute of Technology Madras

Design of Steel Structures

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If the beam is non-prismatic within the lateral supports and has reduced width of
flange at lesser moment section the lateral buckling strength decreases.
The effect of residual stresses is to reduce the lateral buckling capacity. If the
compression flange is wider than tension flange lateral buckling strength increases and
if the tension flange is wider than compression flange, lateral buckling strength
decreases. The residual stresses and hence its effect is more in welded beams as
compared to that of rolled beams.

The initial imperfections in geometry tend to reduce the load carrying capacity.

Fig 6.5 Beam subjected to Non-uniform moment

The design buckling (Bending) resistance moment of laterally unsupported
beams are calculated as per Section 8.2.2 of the code.
If the non-dimensional slenderness λLT ≤ 0.4, no allowance for lateral-torsional buckling
is necessary. Appendix F of the code gives the method of calculating Mcr,, the elastic
lateral torsional buckling moment for difficult beam sections, considering loading and a
support condition as well as for non-prismatic members.

6.3.3 Effective length of compression flanges
The lateral restraints provided by the simply supported condition assumption in
the basic case, is the lowest and therefore the Mcr is also the lowest. It is possible, by
other restraint conditions, to obtain higher values of Mcr, for the same structural section,
which would result in better utilisation of the section and thus, saving in weight of

Indian Institute of Technology Madras

Design of Steel Structures

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material. As lateral buckling involves three kinds of deformation, namely, lateral
bending, twisting and warping, it is feasible to think of various types of end conditions.
But the supports should either completely prevent or offer no resistance to each type of
deformation. Solutions for partial restraint conditions are complicated. The effect of
various types of support conditions is taken into account by way of a parameter called
effective length.

For the beam with simply supported end conditions and no intermediate lateral
restraint the effective length is equal to the actual length between the supports, when a
greater amount of lateral and torsional restraints is provided at support. When the
effective length is less than the actual length and alternatively the length becomes more
when there is less restraint. The effective length factor would indirectly account for the
increased lateral and torsional rigidities by the restraints.

6.3.4 Shear
Let us take the case of an ‘I’ beam subjected to the maximum shear force (at the
support of a simply supported beam). The external shear ‘V’ varies along the
longitudinal axis ‘x’ of the beam with bending moment as V=dM/dx . While the beam is
in the elastic stage, the internal shear stresses τ , which resist the external shear, V, can
be written as,

τ=

VQ
lt

where
V = shear force at the section
I = moment of inertia of the entire cross section about the neutral axis
Q = moment about neutral axis of the area that is beyond the fibre at which τ is
calculated and ‘t’ is the thickness of the portion at which τ is calculated.

Indian Institute of Technology Madras

Design of Steel Structures

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The above Equation is plotted in Fig. 6.6, which represents shear stresses in the
elastic range. It is seen from the figure that the web carries a significant proportion of
shear force and the shear stress distribution over the web area is nearly uniform.
Hence, for the purpose of design, we can assume without much error that the average
shear stress as

τav =

V
tw dw

where
tw = thickness of the web
dw = depth of the web
The nominal shear yielding strength of webs is based on the Von Mises yield
criterion, which states that for an un-reinforced web of a beam, whose width to
thickness ratio is comparatively small (so that web-buckling failure is avoided), the
shear strength may be taken as

τy =

fy
3

= 0.58f y

where
fy = yield stress.
The shear capacity of rolled beams Vc can be calculated as
Vc ≈ 0.6 f y t w d w

Fig 6.6 Elastic shear stresses

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Design of Steel Structures

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When the shear capacity of the beam is exceeded, the ‘shear failure’ occurs by
excessive shear yielding of the gross area of the webs as shown in Fig 6.7. Shear
yielding is very rare in rolled steel beams.

Fig 6.7 Shear yielding near support
The factored design shear force V in the beam should be less than the design
shear strength of web. The shear area of different sections and different axes of
bending are given in Section 8.4.1.1

6.3.4.1 Resistance to shear buckling

Fig 6.8 Buckling of a girder web in shear

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Fig 6.9 A typical plate girder
The girder webs will normally be subjected to some combination of shear and
bending stresses. The most severe condition in terms of web buckling is normally the
pure shear case. It follows that it is those regions adjacent to supports or the vicinity of
point loads, which generally control the design. Shear buckling occurs largely as a result
of the compressive stresses acting diagonally within the web, as shown in Fig.6. 8 with
the number of waves tending to increase with an increase in the panel aspect ratio c / d.

When d /tw≤ 67ε where ε = (250 / fy)0.5 the web plate will not buckle because the
shear stress τ is less than critical buckling stress ‘τcr’. The design in such cases is
similar to the rolled beams here. Consider plate girders having thin webs with d/tw > 67ε.
In the design of these webs, shear buckling should be considered. In a general way, we
may have an un-stiffened web, a web stiffened by transverse stiffeners (Fig. 6.9) and a
web stiffened by both transverse and longitudinal stiffeners (Fig. 6.10)

Fig 6.10 End panel strengthened by longitudinal stiffener

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Design of Steel Structures

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6.3.4.2 Shear buckling design methods
The webs, designed either with or without stiffeners and governed by buckling
may be evaluated by using two methods

1.

Simple Post Critical Method

2.

Tension Field Method

6.3.4.2.1 Simple post critical method
It is a simplified version of a method for calculating the post-buckled member
stress. The web possesses considerable post-buckling strength reserve and is shown in
Fig. 6.11.

When a web plate is subjected to shear, we can visualize the structural
behaviour by considering the effect of complementary shear stresses generating
diagonal tension and diagonal compression. Consider an element E in equilibrium
inside a square web plate with shear stress q. The requirements of equilibrium result in
the generation of complementary shear stresses as shown in Fig.6.12. This result in the
element being subjected to principal compression along the direction AC and tension
along the direction BD. As the applied loading is incrementally enhanced, with
corresponding increases in q, very soon, the plate will buckle along the direction of
compression diagonal AC.

The plate will lose its capacity to any further increase in compressive stress. The
corresponding shear stress in the plate is the "critical shear stress" τcr. The value of τcr
can be determined from classical stability theory, if the boundary conditions of the plate
are known. As the true boundary conditions of the plate girder web are difficult to
establish due to restraints offered by flanges and stiffeners we may conservatively
assume them to be simply supported. The critical shear stress in such a case is given
by

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τcr =

(

k r π2 E

12 1 = µ 2

) ( d / t w )2

Fig. 6.11 Postbuckling reserve strength of web

kv = 5.35 When the transverse stiffeners is provided at the support,

kv=4.0 +5.35(c/d)2 for c/d < 1.0 i.e., for webs with closely spaced transverse stiffeners.

kv=5.35 +4(c/d)2 for c/d ≥1.0 for wide panels.

Fig 6.12 Unbuckled shear panel

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When the value of (d/t) is sufficiently low (d/t < 85) τcr increases above the value
of yield shear stress, and the web will yield under shear before buckling.
Based on this theory, the code gives the following values for τcr for webs, which are not
too slender (Section 8.4.2.2a). The values depend on the slenderness parameter λw as
defined in the code.

6.3.4.2.2. Tension field methods
Design of plate girders with intermediate stiffeners, as indicated in Fig. 6.10, can
be done by limiting their shear capacity to shear buckling strength. However, this
approach is uneconomical, as it does not account for the mobilisation of the additional
shear capacity as indicated earlier. The shear resistance is improved in the following
ways:
i.
ii.

Increasing in buckling resistance due to reduced c/d ratio;
The web develops tension field action and this resists considerably larger

stress than the elastic critical strength of web in shear

Figure 6.13 shows the diagonal tension fields anchored between top and bottom
flanges and against transverse stiffeners on either side of the panel with the stiffeners
acting as struts and the tension field acting as ties. The plate girder behaves similar to
an N-truss Fig.6.14.
The nominal shear strength for webs with intermediate stiffeners can be
calculated by this method according to the design provision given in code.

Fig 6.13 Tension field in individual sub-panel

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Fig 6.14 Tension field action and the equivalent N-truss

6.3.5 Stiffened web panels
For tension field action to develop in the end panels, adequate anchorage should be
provided all around the end panel. The anchor force Hq required to anchor the tension
field force is

⎛
V
H q = 1.5Vdp ⎜1 − cr
⎜ Vdp
⎝

1/ 2

⎞
⎟
⎟
⎠

The end panel, when designed for tension field will impose additional loads on
end post; hence, it will become stout (Fig 8.2 of the code). For a simple design, it may
be assumed that the capacity of the end panel is restricted to Vcr, so that no tension
field develops in it (Fig 8.1 of the code). In this case, end panel acts as a beam
spanning between the flanges to resist shear and moment caused by Hq and produced
by tension field of penultimate panel.

This approach is conservative, as it does not utilise the post-buckling strength of
end panel especially where the shear is maximum. This will result in c/d value of the
end panel spacing to be less than of other panels. The end stiffeners should be
designed for compressive forces due to bearing and the moment Mtf, due to tension field
in the penultimate panel in order to be economical the end panel also may be designed
using tension field action. In this case, the bearing stiffeners and end post are designed

Indian Institute of Technology Madras

Design of Steel Structures

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for a combination due to bearing and a moment equal to 2/3 caused due to tension in
the flange Mtf, instead of one stout stiffener we can use a double stiffener as shown in
Fig. 8.3 of the code. Here, the end post is designed for horizontal shear and moment
Mtf.

6.3.6 Design of beams and plate girders with solid webs.
The high bending moment and shear forces caused by carrying heavy loads over
long spans may exceed the capacity of rolled beam sections. Plate girders can be used
in such cases and their proportions can be designed to achieve high strength/weight
ratio. In a plate girder, it can be assumed that the flanges resist the bending moment
and the web provides resistance to the shear force. For economic design, low flange
size and deep webs are provided. This results in webs for which shear failure mode is a
consequence of buckling rather than yielding.

Minimum web thickness − In general, we may have unstiffened web, a web
stiffened by transverse stiffeners, (Fig 6.9) or web stiffened by both transverse and
longitudinal stiffeners (Fig 6.10).

By choosing a minimum web thickness tw, the self-weight is reduced. However,
the webs are vulnerable to buckling and hence, stiffened if necessary. The web
thickness based on serviceability requirement is recommended in Section 8.6.1.1 of the
code.

6.3.6.1 Compression flange buckling requirement
Generally, the thickness of flange plate is not varied along the span of plate
girders. For non-composite plate girder the width of flange plates is chosen to be about
0.3 times the depth of the section as a thumb rule. It is also necessary to choose the
breadth to thickness ratio of the flange such that the section classification is generally
limited to plastic or compact section only. This is to avoid local buckling before reaching

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the yield stress. In order to avoid buckling of the compression flange into the web, the
web thickness shall be based on recommendation given in Section 8.6.1.2 of the code.

6.3.6.2 Flanges
For a plate girder subjected to external loading the minimum bending moment
occurs at one section usually, e.g. when the plate girder is simply supported at the ends
and subjected to the uniformly distributed load, then, maximum bending moment occurs
at the centre. Since the values of bending moment decreases towards the end, the
flange area designed to resist the maximum bending moment is not required to other
sections. Therefore the flange plate may be curtailed at a distance from the centre of
span greater than the distance where the plate is no longer required as the bending
moment decreases towards the ends.

Usually, two flange angles at the top and two flange angles at the bottom are
provided. These angles extend from one end to the other end of the girder. For a good
proportioning, the flange angles must provide an area at least one-third of the total
flange area.

Generally, horizontal flange plates are provided to and connected to the
outstanding legs of the flange angles. The flange plates provide an additional width to
the flange and thus, reduced the tendency of the compression flange to buckle. These
plates also contribute considerable moment of inertia for the section of the girder it gives
economy as regards the material and cost. At least one flange plate should be run for
the entire length of the girder

6.3.6.3 Flange splices
A joint in the flange element provided to increase the length of flange plates is
known as flange splice. The flange plate should be avoided as far as possible.
Generally, the flange plates can be obtained for full length of the plate girder. In spite of

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the availability of full length of flange plates, sometimes, it becomes necessary to make
flange splices. Flange joints should not be located at the points of main bending
moment. The design provisions for flange splices are given in detail in Section 8.6.3.2 of
the code.

6.3.6.3.1 Connection of flange to web
Rivets/bolts or weld connecting the flanges angles and the web will be subjected
to horizontal shear and sometimes vertical loads which may be applied directly to the
flanges for the different cases such as depending upon the directly applied load to the
either web or flange and with or without consideration of resistance of the web.

6.3.6.3.2 Splices in the web
A joint in the web plate provided to increase its length is known as web splices.
The plates are manufactured upto a limited length. When the maximum manufactured
length of the plate is insufficient for full length of the plate girder web splice becomes
essential, when the length of plate girder is too long to handle conveniently during
transportation and erection. Generally, web splices are not used in buildings; they are
mainly used in bridges.

Splices in the web of the plate girder are designed to resist the shear and
moment at the spliced section. The splice plates are provided on each side of the web.
Groove-welded splice in plate girders develop the full strength of the smaller spliced
section. Other types of splices in cross section of plate girders shall develop the
strength required by the forces at the point of the splice.

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6.3.7 Stiffener design
6.3.7.1 General
These are members provided to protect the web against buckling. The thin but
deep web plate is liable to vertical as well as diagonal buckling. The web may be
stiffened with vertical as well as longitudinal stiffeners

Stiffeners may be classified as
a)

Intermediate transverse web stiffeners

b)

Load carrying stiffeners

c)

Bearing stiffeners

d)

Torsion stiffeners

e)

Diagonal stiffeners and

f)

Tension stiffeners

The functions of the different types of stiffeners are explained in the Section 8.7.1.1 of
the code.

6.3.7.2 Stiff bearing length
The application of heavy concentrated loads to a girder will produce a region of
very high stresses in the part of the web directly under the load. One possible effect of
this is to cause outwards buckling of this region as if it were a vertical strut with its ends
restrained by the beam's flanges. This situation also exists at the supports where the
'load' is now the reaction and the problem is effectively turned upside down. It is usual
to interpose a plate between the point load and the beam flange, whereas in the case of
reactions acting through a flange, this normally implies the presence of a seating cleat.

Indian Institute of Technology Madras

Design of Steel Structures

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Fig 6.15 Dispersion of concentrated loads and reactions
In both cases, therefore, the load is actually spread out over a finite area by the
time it passes into the web as shown in Fig.6.15. It is controlled largely by the
dimensions of the plate used to transfer the load, which is itself termed "the stiff length
of bearing"

Outstand of web stiffeners and eccentricity of the stiffeners is explained as per
Section 8.7.1.2 to 8.7.1.5 of the code.

6.3.7.3 Design of Intermediate transverse web stiffeners
Intermediate transverse stiffeners are provided to prevent out of plane buckling of
web at the location of the stiffeners due to the combined effect of bending moment and
shear force.

Intermediate transverse stiffeners must be proportioned so as to satisfy two
conditions

1)

They must be sufficiently stiff not to deform appreciably as the web tends to

buckle.
2)

They must be sufficiently strong to withstand the shear transmitted by the

web.

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Since it is quite common to use the same stiffeners for more than one task (for
example, the stiffeners provided to increase shear buckling capacity can also be
carrying heavy point loads), the above conditions must also, in such cases, include the
effect of additional direct loading.

The condition (1) is covered by Section 8.7.2.4 of the code.

The strength requirement is checked by ensuring that the stiffeners acting as a
strut is capable of withstanding Fq. The buckling resistance Fq of the stiffeners acting as
strut (with a cruciform section as described earlier) should not be less than the
difference between the shear actually present adjacent to the stiffeners V, and the shear
capacity of the (unstiffened) web Vcr together with any coexisting reaction or moment.
Since the portion of the web immediately adjacent to the stiffeners tends to act with it,
this "strut" is assumed to consist also of a length of web of 20 t on either side of the
stiffeners centre line giving an effective section in the shape of a cruciform. Full details
of this strength check are given in Section 8.7.2.5 of the code. If tension field action is
being utilized, then the stiffeners bounding the end panel must also be capable of
accepting the additional forces associated with anchoring the tension field.

6.3.7.4 Load carrying stiffeners

Fig.6.16 Cruciform section of the load carrying stiffener

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Whenever there is a risk of the buckling resistance of the web being exceeded,
especially owing to concentrated loads, load-carrying stiffeners are provided. The
design of load-carrying stiffeners is essentially the same as the design of vertical
stiffeners. The load is again assumed to be resisted by strut comprising of the actual
stiffeners plus a length of web of 20t on either side, giving an effective cruciform section.
Providing the loaded flange is laterally restrained, the effective length of the 'strut' may
be taken as 0.7L. Although no separate stiffener check is necessary, a load-bearing
stiffener must be of sufficient size that if the full load were to be applied to them acting
independently, i.e., on a cross-section consisting of just the stiffeners, as per Fig 6.16.
Then the stress induced should not exceed the design strength by more than 25%. The
bearing stress in the stiffeners is checked using the area of that portion of the stiffeners
in contact with the flange through which compressive force is transmitted.

6.3.7.5 Bearing stiffeners
Bearing stiffeners are required whenever concentrated loads, which could cause
vertical buckling of web of the girder, are applied to either flange. Such situations occur
on the bottom flange at the reactions and on the top flange at the point of concentrated
loads. Figure 6.17 shows bearing stiffeners consisting of plates welded to the web. They
must fit tightly against the loaded flange. There must be sufficient area of contact
between the stiffeners and the flange to deliver; the load without exceeding the
permissible bearing on either the flange material or the stiffeners must be adequate
against buckling and the connection to the web must be sufficient to transmit the load as
per Section 8.7.4 of the code.

Fig. 6.17 Bearing stiffeners

Indian Institute of Technology Madras

Design of Steel Structures

Prof. S.R.Satish Kumar and Prof. A.R.Santha Kumar

The bearing stress on the contact area between stiffeners and flanges is
analogous to the compressive stress at the junction of web and flange of rolled beams
subjected to concentrated load.

Since buckling of bearing stiffeners is analogous to buckling of web at point of
concentrated load, the required moment of inertia is not easy to evaluate. The buckled
stiffeners may take any forms depending on the manner in which the flanges are
restrained. In most cases, the compression flange of the girder will be supported
laterally at points of concentrated load by bracing or by beams framing into it, so
buckling will approximate the form of an end-fixed column. Even if the flanges are free
to rotate, the stiffeners need not be considered as end-hinged columns, because the
load concentrated on one end of the stiffeners is resisted by forces distributed along its
connection to the web instead of by a force concentrated at the opposite end as in
columns.

The connection to the web is merely a matter of providing sufficient welding to
transmit the calculated load on the stiffeners as per the Section 8.7.6 of the code.

Design of different types of stiffeners is given in the code from Sections 8.7.4 to
8.7.9.

6.3.7.6 Connection of web of load carrying and bearing
The web connection of load carrying stiffeners to resist the external load and
reactions through flange shall be designed as per the design criteria given in the code.

6.3.7.7 Horizontal stiffeners
Horizontal stiffeners are generally not provided individually. They are used in
addition to vertical stiffeners, which are provided close to the support to increase the
bearing resistance and to improve the shear capacity.

Indian Institute of Technology Madras

Design of Steel Structures

Prof. S.R.Satish Kumar and Prof. A.R.Santha Kumar

The location and placing of horizontal stiffeners on the web are based on the
location of neutral axis of the girder. Thickness of the web tw and second moment of
area of the stiffeners Is are as per the conditions and design provisions given in Section
8.7.13 of the code.

6.3.8 Box girders
The design and detailing of box girders shall be such as to give full advantage of
its higher load carrying capacity.

Diaphragm shall be used where external vertical as well as transverse forces are
to be transmitted from one member to another. The diaphragms and their fastenings
shall be proportioned to distribute other force applied to them and in addition, to resist
the design transverse force and the resulting shear forces. The design transverse force
shall be taken as shared equally between the diaphragms.

When concentrated loads are carried from one beam to the other or distributed
between the beams, diaphragms having sufficient stiffeners to distribute the load shall
be bolted or welded between the beams. The design of members shall provide for
torsion resulting from any unequal distribution of loads.

6.3.9 Purlin and sheeting rails
Purlins attached to the compression flange of a main member would normally be
acceptable as providing full torsional restraint; where purlins are attached to tension
flange, they should be capable of providing positional restraint to that flange but are
unlikely (due to the rather light purlin/rafter connections normally employed) to be
capable of preventing twist and bending moment based on the lateral instability of the
compression flange.

Indian Institute of Technology Madras

Design of Steel Structures

Prof. S.R.Satish Kumar and Prof. A.R.Santha Kumar

6.3.10 Bending in a non-principal plane
When deflections are constrained to a non-principal plane by the presence of
lateral restraints, the principal axes bending moments are calculated due to the restraint
forces as well as the applied forcs by any rational method. The combined effect is
verified using the provisions in Section 9. Similarly, when the deflections are
unconstrained due to loads acting in a non-principal plane, the principal axes bending
moments are arrived by any rational method. The combined effects have to satisfy the
requirements of Section 9.

Indian Institute of Technology Madras